How do you design a single-plate shear tab connection?

A single-plate shear tab is the most common steel beam-to-column connection. Design per AISC 360-22 checks in sequence: (1) Bolt shear — phi Rn = phi x Fnv x Ab x n. For 3/4 in A325-N bolts (Fnv=54 ksi, Ab=0.442 in^2, phi=0.75): 4 bolts provide phi Rn = 0.75 x 54 x 0.442 x 4 = 71.5 kips. (2) Bolt bearing on plate — phi Rn = phi x 1.2 x Lc x t x Fu for standard holes. Lc = clear distance between bolt holes or from hole to edge. For 3/8 in A36 plate: interior bolt bearing = 44.1 kips/bolt, edge bolt = 22.1 kips. (3) Plate shear yielding — phi Rn = 1.0 x 0.60 x Fy x Agv. For 3/8 x 12 in plate: 97.2 kips. (4) Weld to column — phi Rn = phi x 0.60 x FEXX x (0.707 x w) x L. For 5/16 in fillet E70XX both sides: 167 kips. The governing limit state (lowest capacity) controls. All must exceed the factored reaction Ru.

How is bolt bearing capacity calculated per AISC 360?

Bolt bearing capacity per AISC 360-22 Section J3.10 depends on hole type and deformation consideration: For standard holes WITH deformation considered (typical): phi Rn = phi x 1.2 x Lc x t x Fu <= phi x 2.4 x d x t x Fu. For standard holes WITHOUT deformation considered (oversized holes in slotted connections): phi Rn = phi x 1.0 x Lc x t x Fu <= phi x 2.0 x d x t x Fu. Where: Lc = clear distance between bolt holes = s - dh (spacing minus hole diameter) for interior bolts, or Le - dh/2 (edge distance minus half hole diameter) for edge bolts. d = nominal bolt diameter. t = plate thickness. Fu = plate tensile strength. The Lc-based formula governs when bolt spacing is tight; the 2.4d-based upper bound governs when bolts are widely spaced and bearing is limited by the plate tearing out. Key pitfall: edge bolts often have HALF the capacity of interior bolts because Lc for edge bolts is typically much smaller than for interior bolts.

What are the limit states to check in steel connection design?

Every steel connection must be checked for ALL applicable limit states per AISC 360 Chapter J. The complete sequence: (1) Bolt shear — threads included (N) or excluded (X) from shear plane affects Fnv. (2) Bolt bearing/tearout — at each bolt location, controlled by Lc or 2.4d upper bound. (3) Block shear rupture — a combination of tension and shear on a block of plate material, per AISC J4.3. (4) Plate gross shear yielding — phi=1.0 x 0.60Fy x Agv. (5) Plate net shear rupture — phi=0.75 x 0.60Fu x Anv. (6) Weld strength — phi=0.75 x 0.60 x FEXX x throat area for fillet welds; phi=0.80 for CJP groove welds. (7) Flexural yielding of the plate (for extended shear tabs). The CONNECTION capacity is the MINIMUM of all limit states — not the average, not the bolt capacity alone. A connection that passes bolt shear but fails block shear is not adequate. This is the most common error in connection design: checking only bolts and neglecting the plate limit states.

How is fillet weld strength calculated per AISC 360?

Fillet weld design strength per AISC 360-22 Section J2.4: phi Rn = phi x 0.60 x FEXX x (0.707 x w) x L. Where: phi = 0.75 for fillet welds. FEXX = electrode classification strength (70 ksi for E70XX electrodes). 0.707 x w = effective throat thickness (w = leg size). L = total weld length. For a 5/16 in fillet weld: throat = 0.707 x 5/16 = 0.221 in. Strength per inch = 0.75 x 0.60 x 70 x 0.221 = 6.97 kips/inch. For a 12 in weld on each side (L=24 in total): phi Rn = 6.97 x 24 = 167 kips. Minimum fillet weld size per AISC Table J2.4: 3/16 in for 1/4 in plate, 1/4 in for 1/2 in plate, 5/16 in for 3/4 in plate. Maximum fillet weld size: for plate = 1/4 in, w <= plate thickness minus 1/16 in. The minimum size prevents cracking from rapid cooling; the maximum prevents excessive melting of the base metal edge.

What are the key differences between shear and moment connections?

Shear connections (simple connections) transfer only shear — they allow end rotation and are modeled as pins in analysis. Types: single-plate shear tabs, double angles, end plates with short projection, and seated connections. Design checks: bolt shear, bearing, block shear, plate shear, weld. Moment connections (rigid/full-moment) transfer both shear AND moment — they restrain end rotation. Types: extended end plates, flange plates, and directly welded flanges. Additional design checks beyond shear: flange force = Mu/(d - tf), bolt tension in the tension flange region (including prying action per AISC 360 Section J3.7), end plate flexural yielding (yield line analysis), column flange bending, column web panel zone shear, and column web local yielding and crippling. Moment connections require 5-10x more design checks than shear connections because they must transfer flange forces that can be 100-300 kips each. The connection is often the governing element in moment frame design, not the beam or column.

Steel Connection Design Examples -- Four Worked Problems per AISC 360-22

Q: What are the limit states to check in steel connection design?

Every steel connection must be checked for ALL applicable limit states per AISC 360 Chapter J. The complete sequence: (1) Bolt shear — threads included (N) or excluded (X) from shear plane affects Fnv. (2) Bolt bearing/tearout — at each bolt location, controlled by Lc or 2.4d upper bound. (3) Block shear rupture — a combination of tension and shear on a block of plate material, per AISC J4.3. (4) Plate gross shear yielding — phi=1.0 x 0.60Fy x Agv. (5) Plate net shear rupture — phi=0.75 x 0.60Fu x Anv. (6) Weld strength — phi=0.75 x 0.60 x FEXX x throat area for fillet welds; phi=0.80 for CJP groove welds. (7) Flexural yielding of the plate (for extended shear tabs). The CONNECTION capacity is the MINIMUM of all limit states — not the average, not the bolt capacity alone. A connection that passes bolt shear but fails block shear is not adequate. This is the most common error in connection design: checking only bolts and neglecting the plate limit states.

Q: How is fillet weld strength calculated per AISC 360?

Fillet weld design strength per AISC 360-22 Section J2.4: phi Rn = phi x 0.60 x FEXX x (0.707 x w) x L. Where: phi = 0.75 for fillet welds. FEXX = electrode classification strength (70 ksi for E70XX electrodes). 0.707 x w = effective throat thickness (w = leg size). L = total weld length. For a 5/16 in fillet weld: throat = 0.707 x 5/16 = 0.221 in. Strength per inch = 0.75 x 0.60 x 70 x 0.221 = 6.97 kips/inch. For a 12 in weld on each side (L=24 in total): phi Rn = 6.97 x 24 = 167 kips. Minimum fillet weld size per AISC Table J2.4: 3/16 in for 1/4 in plate, 1/4 in for 1/2 in plate, 5/16 in for 3/4 in plate. Maximum fillet weld size: for plate = 1/4 in, w <= plate thickness minus 1/16 in. The minimum size prevents cracking from rapid cooling; the maximum prevents excessive melting of the base metal edge.

Q: What are the key differences between shear and moment connections?

Shear connections (simple connections) transfer only shear — they allow end rotation and are modeled as pins in analysis. Types: single-plate shear tabs, double angles, end plates with short projection, and seated connections. Design checks: bolt shear, bearing, block shear, plate shear, weld. Moment connections (rigid/full-moment) transfer both shear AND moment — they restrain end rotation. Types: extended end plates, flange plates, and directly welded flanges. Additional design checks beyond shear: flange force = Mu/(d - tf), bolt tension in the tension flange region (including prying action per AISC 360 Section J3.7), end plate flexural yielding (yield line analysis), column flange bending, column web panel zone shear, and column web local yielding and crippling. Moment connections require 5-10x more design checks than shear connections because they must transfer flange forces that can be 100-300 kips each. The connection is often the governing element in moment frame design, not the beam or column.

This page presents four complete worked examples of common steel connections. Each example follows AISC 360-22 LRFD provisions and shows every calculation step explicitly. The connections are independent -- you can study them in any order.

For the underlying limit state equations and bolt/weld capacity formulas, see the Steel Connection Design Guide. For a catalog of connection types and selection guidance, see Connection Types Explained.

Example 1: Double-Angle Shear Connection (Beam-to-Girder)

Problem: Connect a W16x31 floor beam to the web of a W24x76 girder. Vu = 42 kip (LRFD). The beam is coped at the top flange only. Use A36 angles and 3/4 in. A325-N bolts in standard holes (bearing-type connection).

Beam geometry: W16x31: d = 15.9 in., bf = 5.53 in., tw = 0.275 in. Cope: 1-1/2 in. deep x 3 in. long at top flange.

Step 1 -- Select angles and bolt layout: Use 2L 4 x 3 x 5/16 (long leg to beam web, short leg outstanding). Two rows of 3 bolts through the beam web, two rows of 3 bolts through the girder web. Total 6 bolts in the beam web, 6 bolts in the girder web.

Step 2 -- Beam web bolts (3 per vertical row): Single shear per bolt: phi Rn = 0.75 x 54 x 0.442 = 17.9 kip. Eccentricity = 2.25 in. (distance from bolt line to angle heel). From AISC Manual Table 7-6 for 3 bolts at 3 in. spacing, e = 2.25 in.: C = 3.99. phi Rn_group = 0.75 x 3.99 x 17.9 = 53.6 kip > 42 kip. OK.

Step 3 -- Angle shear rupture (outstanding leg, girder side): The angles transfer the full shear. Shear area per angle = L_angle x t = 8.5 x 0.3125 = 2.66 in^2. Two angles: Agv = 5.31 in^2. phi Rn = 0.90 x 0.60 x 36 x 5.31 = 103 kip >> 42 kip. OK.

Step 4 -- Angle bearing on beam web: Bearing on 5/16 in. angle (Fu = 58 ksi). Lc = 3 - 13/16 = 2.1875 in. Tearout: 1.2 x 2.1875 x 0.3125 x 58 = 47.6 kip per bolt. Bearing: 2.4 x 0.75 x 0.3125 x 58 = 32.6 kip per bolt. Bearing controls. phi Rn = 0.75 x 32.6 = 24.5 kip per bolt. 6 bolts: 147 kip >> 42 kip.

Step 5 -- Block shear in beam web (coped region): The cope reduces the effective shear area. Remaining web depth = 15.9 - 1.5 (cope) = 14.4 in. However, the bolts only engage the top 8.5 in. of this. Effective tension area Ant = (1.5 + 2x3 - 2.5x13/16) x 0.275 = (7.5 - 2.03) x 0.275 = 1.50 in^2. Shear area Anv = (8.5 - 2.5x13/16) x 0.275 = (8.5 - 2.03) x 0.275 = 1.78 in^2. Agv = 8.5 x 0.275 = 2.34 in^2.

Rn = 0.60 x 65 x 1.78 + 1.0 x 65 x 1.50 = 69.4 + 97.5 = 166.9 kip. Cap: 0.60 x 50 x 2.34 + 1.0 x 65 x 1.50 = 70.2 + 97.5 = 167.7 kip. phi Rn = 0.75 x 166.9 = 125 kip >> 42 kip. OK.

Step 6 -- Girder web bearing and shear: The girder web (tw = 0.440 in. for W24x76) is thicker than the angle, so bearing on the angle controls. Web shear in the girder at the connection must be checked per AISC 360 Section G2, but for a typical interior connection this is rarely controlling.

Final design: 2L 4 x 3 x 5/16 x 0'-8-1/2 (A36). Six 3/4 in. A325-N bolts through the beam web at 3 in. c/c, six 3/4 in. A325-N bolts through the girder web at 3 in. c/c.

Example 2: Extended Shear Tab (Beam-to-Column Flange)

Problem: Connect a W21x50 floor beam to the flange of a W14x90 column. Vu = 85 kip. Use a single-plate shear tab. The beam is coped top and bottom. A36 plate. 3/4 in. A325-N bolts. E70XX electrodes.

Beam geometry: W21x50: d = 20.8 in., tw = 0.380 in., bf = 6.53 in.

Step 1 -- Trial plate and bolt count: Try 5 bolts at 3 in. spacing. Plate depth = (5-1) x 3 + 2 x 1.5 = 15 in. For 5 bolts at 3 in. c/c, e = distance from bolt line to weld = a/2 = 3 in. standard.

Step 2 -- Eccentric shear check (IC method): From AISC Manual Table 7-6: for 5 bolts, e = 3 in., s = 3 in., C = 5.60. Single-bolt shear: phi Rn = 0.75 x 54 x 0.442 = 17.9 kip. phi Rn_group = 0.75 x 5.60 x 17.9 = 75.2 kip < 85 kip. Insufficient.

Try 6 bolts. Plate depth = (6-1) x 3 + 2 x 1.5 = 18 in. For 6 bolts, e = 3 in., s = 3 in.: C = 6.97. phi Rn_group = 0.75 x 6.97 x 17.9 = 93.5 kip > 85 kip. OK.

Step 3 -- Plate shear yielding: Ag = 0.375 x 18 = 6.75 in^2. phi Rn = 0.90 x 0.60 x 36 x 6.75 = 131.2 kip > 85 kip. OK.

Step 4 -- Plate shear rupture (net section): Net depth = 18 - 6 x (0.8125 + 0.0625) = 18 - 5.25 = 12.75 in. Anv = 12.75 x 0.375 = 4.78 in^2. phi Rn = 0.75 x 0.60 x 58 x 4.78 = 124.8 kip > 85 kip. OK.

Step 5 -- Bolt bearing on plate: Lc per bolt = 3 - 0.8125 = 2.1875 in. Tearout: 1.2 x 2.1875 x 0.375 x 58 = 57.1 kip. Bearing: 2.4 x 0.75 x 0.375 x 58 = 39.2 kip. Bearing controls. phi Rn = 0.75 x 39.2 = 29.4 kip per bolt. 6 bolts: 176 kip >> 85 kip. OK.

Step 6 -- Block shear: Ant = (1.5 + 5x3 - 5.5x0.8125) x 0.375 = (16.5 - 4.47) x 0.375 = 4.51 in^2. Anv = (18 - 5.5x0.8125) x 0.375 = (18 - 4.47) x 0.375 = 5.07 in^2. Agv = 18 x 0.375 = 6.75 in^2.

Rn = 0.60 x 58 x 5.07 + 1.0 x 58 x 4.51 = 176 + 262 = 438 kip. Cap: 0.60 x 36 x 6.75 + 1.0 x 58 x 4.51 = 146 + 262 = 408 kip. phi Rn = 0.75 x 408 = 306 kip >> 85 kip. OK.

Step 7 -- Fillet weld to column flange: 1/4 in. fillet both sides of plate (18 in. long). phi Rn = 2 x 18 x 5.57 = 200.5 kip >> 85 kip. Including eccentricity on weld group: M = 85 x 3 = 255 kip-in. Weld section modulus: Sx = 2 x 18^2 / 6 = 108 in^2 x in/in. Peak demand = Vu/(2L) + M/(Sx) = 85/36 + 255/108 = 2.36 + 2.36 = 4.72 kip/in. Capacity per 1/4 in. fillet = 5.57 kip/in > 4.72 kip/in. OK.

Step 8 -- Beam web shear (coped): After coping, remaining web depth = 20.8 - 1.5(top) - 1.5(bottom) = 17.8 in. phi Vn = 0.90 x 0.60 x 50 x 17.8 x 0.380 = 183 kip > 85 kip. OK.

Final design: PL 3/8 x 4-1/2 x 1'-6. Six 3/4 in. A325-N bolts at 3 in. c/c. 1/4 in. fillet weld both sides of plate to column flange.

Example 3: Flush End Plate Shear Connection

Problem: Connect a W12x26 beam to the web of a W14x61 column using a flush end plate. Vu = 28 kip. A36 plate, 3/4 in. A325-N bolts.

Beam geometry: W12x26: d = 12.2 in., bf = 6.49 in., tw = 0.230 in.

Step 1 -- Plate geometry: The flush end plate is shop-welded to the beam end and field-bolted to the column. Plate width = bf = 6.5 in. Plate height = d + 1 in. = 13.2 in. Use PL 3/8 x 6-1/2 x 0'-13-1/4.

Step 2 -- Bolt layout: Four 3/4 in. bolts in a single column at 3 in. spacing. Eccentricity = distance from bolt line to weld line = gage/2 = 2.5 in. For 4 bolts, s = 3 in., e = 2.5 in.: from AISC Manual Table 7-6, C = 4.64.

Step 3 -- Bolt group check: phi Rn_group = 0.75 x 4.64 x 17.9 = 62.3 kip > 28 kip. OK.

Step 4 -- End plate shear yielding: Ag = 13.25 x 0.375 = 4.97 in^2. phi Rn = 0.90 x 0.60 x 36 x 4.97 = 96.6 kip > 28 kip. OK.

Step 5 -- Weld (plate to beam web): The shear is transferred from the beam web into the end plate through two vertical fillet welds. Weld length = 12.2 in. (beam depth, minus cope if applicable). Use 3/16 in. fillet. phi Rn = 2 x 12.2 x 4.18 = 102 kip >> 28 kip. Minimum fillet for 0.375 in. plate is 3/16 in. OK.

Final design: PL 3/8 x 6-1/2 x 1'-1-1/4 end plate shop-welded to W12x26 (3/16 in. fillet weld both sides of web to plate). Four 3/4 in. A325-N bolts at 3 in. c/c to column. 13/16 in. diameter standard holes.

Example 4: Column Splice (Partial Penetration Weld)

Problem: Splice a W14x132 upper column to a W14x176 lower column at the 4th floor of a 12-story office tower. Pu = 680 kip (compression), Mu = 180 kip-ft (from wind, minor axis). Use A36 splice plates and PJP groove welds.

Column data: Upper W14x132: d = 14.7 in., bf = 14.7 in., tf = 1.030 in., tw = 0.645 in. Lower W14x176: d = 15.2 in., bf = 15.7 in., tf = 1.310 in.

The lower column is 0.5 in. deeper and 1.0 in. wider. Filler plates are required under the flange splice plates to build out the upper column to the lower column flange width.

Step 1 -- Flange splice plates: The axial compression is transferred primarily through the flanges bearing on each other in direct contact. AISC 360 Section J1.4 requires that column splices in compression have the flanges prepared for full bearing. The flange splice plates provide continuity for moment and tension from uplift.

Flange force (from moment): Tf = Mu / (d - tf) = 180 x 12 / (14.7 - 1.03) = 2,160 / 13.67 = 158 kip per flange.

Flange splice plate area required: Ag_min = 158 / (0.90 x 36) = 4.88 in^2. Use PL 5/8 x 14 on each flange face (Ag = 8.75 in^2 > 4.88 in^2). OK.

Step 2 -- Web splice plate: The web splice transfers the shear (from lateral load eccentricity) and any axial load imbalance. Vu = 25 kip (estimated from frame analysis).

Use PL 3/8 x 10 on each side of the web. Bolt to both upper and lower columns. 4 bolts each side. For 4 bolts, e = 2 in., s = 3 in.: C = 4.60. phi Rn_group = 0.75 x 4.60 x 17.9 = 61.8 kip > 25 kip. OK.

Step 3 -- Weld design: Use PJP groove welds at the flange splice plates, with effective throat = 5/8 in. (matching the plate thickness, but partial penetration). For E70XX electrodes: phi Rn = 0.75 x 0.60 x 70 x (5/8) = 19.7 ksi x effective area. Weld length = 14 in. per flange splice plate. Capacity = 2 x 14 x (5/8) x 0.75 x 0.60 x 70 = 2 x 14 x 0.625 x 31.5 = 551 kip >> 158 kip. OK. Minimum PJP weld size per AWS D1.1 Table 5.1 for 1 in. thick plate is 5/8 in. (PJP, bevel groove).

Step 4 -- Fit-up and erection: The upper column must be milled to bear on the lower column. AISC Code of Standard Practice Section 6.4 requires column ends to be prepared for full bearing within a tolerance of 1/16 in. gap (after erection bolting). The splice plates bolt the sections together during erection. After final plumbing, the connection bolts are tightened.

Final design:

Flange splice plates: 2 PL 5/8 x 14 x 1'-2 (outer plates, one each side of each flange)
Filler plates: PL 1/4 x 14 x 1'-2 inside outer plates at lower column (compensates for thickness difference)
Web splice plates: 2 PL 3/8 x 10 x 1'-2
Bolts: 3/4 in. A325-N, 6 per flange plate (2 rows of 3), 4 per web plate (2 rows of 2)
Welds: PJP bevel groove, 5/8 in. effective throat at flange plates. 1/4 in. fillet at web plates at mill.

For more examples and limit state derivations, see the Steel Connection Design Guide.

Related Tools and References

Disclaimer

This page is for educational and reference use only. The worked examples illustrate typical design approaches but do not replace project-specific engineering analysis. All connection designs must be independently verified by a licensed Professional Engineer (PE) or Structural Engineer (SE). The site operator disclaims liability for any loss arising from the use of this information. Results are PRELIMINARY -- NOT FOR CONSTRUCTION.